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Seismic performance of a 3D full‐scale high‐
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resisting structure—Part I: Design...
Article in Earthquake Engineering & Structural Dynamics · November 2008
DOI: 10.1002/eqe.829
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EARTHQUAKE ENGINEERING AND STRUCTURAL DYNAMICS
Earthquake Engng Struct. Dyn. (2008)
Published online in Wiley InterScience (www.interscience.wiley.com). DOI: 10.1002/eqe.829
Seismic performance of a 3D full-scale high-ductility
steel–concrete composite moment-resisting structure—Part I:
Design and testing procedure
A. Braconi1, ‡ , O. S. Bursi2, § , G. Fabbrocino3, ¶ , W. Salvatore4, ∗, †, and R. Tremblay5, §
1 Corporate
Research Policies of Riva Group S.p.A. ILVA Works, Genova, Italy
of Structural and Mechanical Engineering, University of Trento, Trento, Italy
3 Department of SAVA Engineering & Environment Division, University of Molise, Termoli (CB), Italy
4 Department of Structural Engineering, University of Pisa, Pisa, Italy
5 Group for Research in Structural Engineering, Ecole Polytechnique, Montreal, Canada H3C3A7
2 Department
SUMMARY
A multi-level pseudo-dynamic (PSD) seismic test programme was performed on a full-scale three-bay twostorey steel–concrete composite moment-resisting frame built with partially encased composite columns
and partial-strength connections. The system was designed to provide strength and ductility for earthquake
resistance with energy dissipation located in ductile components of beam-to-column joints including
flexural yielding of beam end-plates and shear yielding of the column web panel zone. In addition, the
response of the frame depending on the column base yielding was analysed. Firstly, the design of the test
structure is presented in the paper, with particular emphasis on the ductile detailing of beam-to-column
joints. Details of the construction of the test structure and the test set-up are also given. The paper then
provides a description of the non-linear static and dynamic analytical studies that were carried out to
preliminary assess the seismic performance of the test structure and establish a comprehensive multi-level
PSD seismic test programme. The resulting test protocol included the application of a spectrum-compatible
earthquake ground motion scaled to four different peak ground acceleration levels to reproduce an elastic
response as well as serviceability, ultimate, and collapse limit state conditions, respectively. Severe damage
to the building was finally induced by a cyclic test with stepwise increasing displacement amplitudes.
Copyright q 2008 John Wiley & Sons, Ltd.
Received 14 March 2007; Revised 16 April 2008; Accepted 1 May 2008
∗ Correspondence
to: W. Salvatore, Department of Structural Engineering, University of Pisa, Pisa, Italy.
E-mail: [email protected]
‡
Research Engineer.
§ Professor.
¶ Associate Professor.
Assistant Professor.
†
Contract/grant sponsor: Joint Research Centre at Ispra
Contract/grant sponsor: Ecole Polytechnique of Montreal; contract/grant number: 19851-2002-09 P1VS3 ISP IT
Contract/grant sponsor: Natural Sciences and Engineering Research Council of Canada
Copyright q
2008 John Wiley & Sons, Ltd.
A. BRACONI ET AL.
KEY WORDS:
seismic design; ductility; steel–concrete composite frames; partial-strength beam-tocolumn joints; composite members; pseudo-dynamic testing
1. INTRODUCTION
Steel–concrete composite moment-resisting frames represent an attractive alternative to their bare
steel counterparts in view of their inherent fire resistance and their relatively higher lateral stiffness
compared with steel only framing constructions. In addition, beam-to-column joints can generally
be obtained at a lower cost in composite structures by taking advantage of the concrete slab to
develop the required flexural stiffness and strength under gravity and lateral loads [1]. However,
common layouts of composite flexural members lead to an asymmetrical response to bending. In
fact, hogging bending can lead to extensive use of reinforcement in the concrete components [2].
Therefore, ductility of members and design requirements for connections can be critical. In addition, with regard to seismic design, beam and column sizes are controlled by lateral drift limits,
so that an interesting design option is represented by partial-strength (PS) beam-to-column joints
detailed to accommodate relevant inelastic rotations through ductile inelastic connection components’ response. These components are designed to bear forces associated with all prescribed load
combinations, taking advantage of the moment redistribution owing to the inelastic connection
response. Global beam hinging mechanisms for earthquake resistance can be achieved at a reduced
cost as the force demand on joints and columns is governed by the expected capacity of the
ductile connection components rather than by the beam flexural capacity. The behaviour of such
a framing system heavily depends on the joint response and a significant part of past research on
steel–concrete composite PS frames was devoted to the development of connection details able
to behave in a ductile manner under cyclic inelastic loading [3–10]. Numerical studies were also
performed to verify the global seismic performance of partially restrained (PR) composite frame
structures [11–13]. This data set is still limited and exhaustive design guidelines to ensure a proper
ductile seismic performance of this system are still missing in current building codes.
Despite their mentioned advantages, PR/PS composite moment-resisting frames were not
commonly used in the past, mainly owing to the lack of effective design procedures available to
practitioners. Information on the seismic performance in terms of ductility and energy dissipation capacity are also lacking, as well as on the constructional and economical characteristics of
the different structural schemes described in the literature, so that the choice of suitable design
solutions is difficult. As an effort to increase the knowledge on the seismic behaviour and design
of PS composite frames, two research projects funded by the European Community, the ECSC
project ‘Applicability of composite structures to sway frames’ [14] and the ECOLEADER project
‘Cyclic and PSD testing of a 3D steel-composite structure’ [15], were recently completed. Their
specific objectives were the analysis of the seismic response of composite frames and the development of design guidelines for high ductility steel–concrete composite structures built with partially
encased composite columns and beam end-plated PS joints of the type illustrated in Figure 1(a).
The main contribution to the energy dissipation in the joints was related to flexural yielding of
beam end-plates and shear yielding of the column web panel zone, which was intentionally not
surrounded by concrete as shown in Figure 1(b). In this context, a two-storey prototype structure
was first designed to obtain realistic member sizes and joint details. Full-scale substructure monotonic and quasi-static cyclic tests on beam-to-column joints were then performed at the University
of Pisa, followed by a pseudo-dynamic (PSD) and a cyclic test programme carried out on a
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
STEEL–CONCRETE COMPOSITE MOMENT-RESISTING STRUCTURE
(a)
Column web panel in shear
(ductile component)
Re-bars in tension
(ductile component)
End-plate and column flange in
bending (ductile component)
Concrete slab in compression
(brittle component)
Bolt in tension
(brittle component)
MRD,PL
MRD,PL
hogging
sagging
Beam flange and web in tension
(ductile component)
Beam flange in compression
(brittle component)
(b)
Column web in compression
(brittle component)
Column web in tension
(ductile component)
Figure 1. Proposed PS composite joint solution: (a) undeformed configuration of a joint and (b) joint
components subdivided as ductile and brittle for capacity design.
full-scale 3D two-storey frame at the European Laboratory for Structural Assessment (ELSA) of
the Joint Research Centre at Ispra, Italy. These tests were complemented with tests on column
base components and joints performed at the University of Naples and at the University of Trento.
The projects also offered the opportunity to examine the viability of the proposed constructional
methods, to calibrate and validate numerical analysis models, and to evaluate the EC8 behaviour
factor [16] for the specific structural system, including assessment of the ductility capacity and
structural overstrength. The present paper deals with the development of the PSD and cyclic test
programme on a full-scale steel–concrete composite-framed building. In detail, the attention is
primarily focussed on the analysis of available design procedures suggested by relevant codes and
on the efficiency of some critical issues of the design process taking into consideration national
and international (basically European and U.S. perspectives) design code backgrounds.
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
A. BRACONI ET AL.
Therefore, a comparative analysis of easy-to-manage design procedures proposed by available
seismic codes versus more refined non-linear response evolution of the structures is issued and
reported herein. Design requirements of the composite framed structure and specific design procedures adopted to implement innovative structural options are reported. Details of the erection phase
and the testing protocol preformed at the ELSA are discussed in order to point out relevant aspects
of the experimental campaign and a detailed report and critical review of experimental results are
reported in a companion paper [17].
2. CURRENT CODE PROVISIONS FOR SEISMIC DESIGN OF COMPOSITE FRAMES
EC8 permits the design of composite frames according to different design concepts based
on different levels of expected yielding (e.g. ductility demand) in structural elements (low
dissipation—ductility class low, DCL; medium dissipation—ductility class medium, DCM; and
high dissipation—ductility class high, DCH) [16]. Different design concepts assign different
dissipative behaviours to the structures. The dissipation induced by plastic deformations can be
located, for DCM and DCH structures, in composite or bare steel parts as beam ends, PR/PS
beam-to-column joint or bracing systems, leading to different dissipative structural types. EC8
does not limit the height of all of these types of structures. Similarly, U.S. AISC seismic codes
[18] suggest three different structural concepts, namely, ordinary, intermediate, and special
moment-resisting composite and bare steel frames, depending on the yielding level, which varies
from very low to high values. AISC provisions consider particular design/detailing rules for
concentrically (C-CBF) and eccentrically braced (C-EBF) and partially restrained composite
frames (C-PRMF) typologies, as the EC8. The ASCE 7-05 code [19] assigns height limitations
to each structural type, differently from EC8; in particular, C-PRMFs are not permitted for high
seismic applications (Seismic Design Categories D and E) and stringent height limits are imposed
for other applications, i.e. 30 m for Seismic Design Category C and 49 m for Seismic Design
Categories A and B. Correspondingly, special steel moment-resisting frames (S-SMRF) have no
height limitation.
2.1. Behaviour factor
Non-linear structural response under seismic loads can be considered elastic by analyses using
a reduced response spectrum if energy dissipation of the structure is ensured through a ductile
behaviour of members and/or other deformation mechanisms. EC8 [16] allows this reduction by
the behaviour factor q; for DCH class steel, the q of composite moment-resisting frames and
PS composite frames is equal to 5u /1 . The ratio u /1 is typically determined by a pushover
analysis and corresponds to the ratio between the lateral loads required to reach the near collapse
condition (global mechanism) and the lateral load at a first significant yielding. A default value of
u /1 = 1.2 is suggested in EC8 [16] for the framed structures; hence, a behaviour factor q equal
to 6.0 is assumed. In the ASCE 7-05 [19], the elastic design spectrum is modified by the response
modification factor R, which is set to 6.0 for C-PRMF and to 8.0 for steel frames. Numerical
studies by Ciutina et al. [12] and Bursi et al. [3] confirm a good performance of multi-storey PS
frames; hence, a design value q = 6.0 was adopted. The particular structure, herein studied, and the
differences between examined codes about height limitations and behaviour factor imply that an
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2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
STEEL–CONCRETE COMPOSITE MOMENT-RESISTING STRUCTURE
experimental assessment of seismic performance of the composite PR-PS MRF should be carried
out to investigate the actual behaviour factor and other structural properties.
2.2. Capacity design
EC8 and AISC provisions adopt ‘capacity design’ procedures for the final design of structural
elements in dissipative seismic-resistant systems. According to this methodology, some structural
elements are chosen and suitably designed to dissipate energy through severe inelastic deformations.
Other elements not devoted to energy dissipation are elastically designed taking into account the
expected capacity of dissipative elements framing to them. Therefore, the ductility demand is
confined in members designed with specific requirements [16, 18]. To develop a global hinging
composite frame mechanism, a strong column–weak beam or PS connection strategy is proposed.
This aim is differently reached by EC8 and AISC. The former imposes that the sum of resisting
bending moments of columns framing into a joint has to be higher than 1.3 times the design values
of resisting moments of beams or design values of resisting moments of PS connections. The latter
imposes that the sum of expected resisting moments of columns has to be higher than the expected
resisting moments of beams or resisting moments of PS connections. In detail, a factor Ry for a
steel grade that takes into account the ratio of the expected yield value to the nominal yield value
must be used, for both columns and beams/connections, and the maximum expected value for
the compressive strength of the concrete is upper limited to 1.2 times f c [18]. In any case the PS
connections shall exhibit a ductile failure mode for energy dissipation accommodating a rotation
capacity consistent with the global deformations expected from the frame. For this reason plastic
deformation must be essentially localized in ductile components of PS/PR composite joints.
2.3. Connection design
In Europe, bolted end-plate beam-to-column joints are commonly used in steel and composite
constructions and component-based design methods are available in Eurocode 3 (EC3) [20] and
Eurocode 4 (EC4) [21] to determine the flexural strength and initial rotational stiffness of connections under monotonic loading. Dissipative semi-rigid and/or PS connections are explicitly referred
in Eurocode 8 (EC8) [16] for moment-resisting frames designed for earthquake resistance, but only
general criteria and behavioural assumptions to be followed are given. Comprehensive detailing
requirements for connection design are not available in EC8 and the design methodologies in EC3
and EC4 do not ensure that a ductile cyclic rotational capacity will be available for PS beam
end-plate joints subjected to seismic loading. For frames of the DCH (high) structural ductility
class, the connection must also exhibit a plastic rotation capability not less than 35 mrad under
cyclic loading without degradation of strength and stiffness greater than 20%. EC8 contains design
requirements for the slab around the columns in moment-resisting frames; however, those were
essentially developed for full-strength connections [22] and were not validated for PS connections.
Such requirements are proposed to ensure the development of two strut-tie mechanisms in the
concrete slab for transferring compressive forces to the column [16]: direct compressive strut,
mechanism (1), and two inclined compressive struts, mechanism (2), Figure 2(d). The U.S. design
criteria for PR/PS composite joints are available for joints consisting of a seat angle, web clip
angles for shear transfer, and a continuous reinforcement in the across column lines for flexural
resistance [23]. In the AISC seismic provisions [18], composite moment frames built with this type
of connection are included in the C-PRMF system category. These frames must be designed so
that yielding mainly occurs in ductile connection components and connection flexibility must be
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
A. BRACONI ET AL.
Seismic steel rebars
Critical Length
Critical Length
Seismic steel rebars
Main
A-A beam
A-A
Main
beam
Secondary
beam
Seismic steel
rebars
Steel Column
(a)
Steel Column
Seismic steel
rebars
(b)
seismic transverse re-bars
(c)
Mechanism 2
Mechanism 1
(d)
Figure 2. Joint details: (a) elevation of an interior joint; (b) elevation of an exterior joint; (c) resistant slab
mechanism according to EC8; and (d) base joint framed on precast foundation blocks.
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
STEEL–CONCRETE COMPOSITE MOMENT-RESISTING STRUCTURE
taken into account in the analysis. The provisions are, however, limited to frames with structural
steel columns, not to composite columns. Selected connections must have a total interstorey drift
capacity of 40 mrad as demonstrated by physical qualification cyclic testing. The AISC recommendation to provide full slab depth at the column was not met in the structure studied herein as
only the portion of the slab located above the steel deck profile was in contact against the column
flange in both the interior and exterior joints [18]. In addition, compressive force transferring from
the concrete slab to the column imposes the installation of transverse reinforcing bars to act as a
tie in the spreading zone of the compressive strut against the column.
2.4. Research significance
Previous research projects have been carried out for assessing the seismic performance of the
steel–concrete composite frames. The ICONS (Intelligent CONtent management Systems) project
deeply investigated the definition of behaviour factor, the effective width of composite beams
in the earthquake-resistant frames, and the design methodology for concrete slab in rigid fullstrength composite beam-to-column joint [24]. In 2004 another research project was carried out
at the University of Taiwan with the aim of assessing the seismic performance of a full-scale
plane composite frame realized with composite columns and steel beams [25, 26]. The final
goal of this research was to examine an innovative pre-cast structural typology and to show the
viability of reinforced concrete-steel beam-to-column connections to provide strength and ductility
to an earthquake-resistant frame. Regardless of the importance and the usefulness of the results
obtained from these previous projects, the knowledge about the seismic performance about PS/PR
composite beam-to-column joint is still limited to experimental results coming from tests on
subassemblages.
Another relevant aspect is related to the study of the seismic design and the inelastic response of
composite column bases. In fact, their details generally fit specifications of bare steel structures even
if the interaction between steel and concrete can activate beneficial effects and energy dissipation.
This is why a test programme performed on a 3D full-scale prototype equipped by PR/PS composite
joints appeared necessary to enhance seismic design rules for this specific structural type until now
not extensively used. As far as column bases are concerned, traditional end-plate connections were
used, but comparative experimental tests with innovative socket-type connections were carried out
at the Laboratory of the University of Naples Federico II in collaboration with the University of
Sannio and the University of Molise [27].
3. DESIGN
3.1. Gravity load resistance
The regular prototype two-storey structure shown in Figure 3(a) was selected to obtain representative dimensions, member sizes, and connection details in order to examine the seismic behaviour
of the structural system. The structure includes five identical two-bay moment-resisting frames
with unequal spans (5+7 m) and spaced 3 m apart. The frames are built with composite beams
connected to partially encased composite columns with PS end-plate joints. Traditional end-plated
connections at the base of the columns were adopted to establish an effective restraint at the
structure/foundation interface. In the direction normal to the moment-resisting frames, lateral resistance is provided by two concentrically braced steel frames located along the exterior walls. Only
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
A. BRACONI ET AL.
(a)
(b)
55
260
150
95
280
WIRE FABRICS Ø6/150
208
280
Composite beam
18
20
107
STEEL SHEETING
Brollo EGB 210
WIRE FABRICS Ø6/150
(c)
189
260
20
150
IPE 300
97
TRANSVERSE REBARS
3Ø16/100cm
91
STUDS NELSON
3/4" x 5"-3/16"
20
TRANSVERSE
REBARS
3Ø16/100cm
18
Composite columns
Figure 3. Prototype structure: (a) moment-resisting frame and designed structure; (b) concrete slab
plan view of the realized prototype at JRC and concentrically braced frame; and (c) main geometrical
features of composite beams and columns.
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
STEEL–CONCRETE COMPOSITE MOMENT-RESISTING STRUCTURE
the response of the structure in the direction of moment-resisting frames depicted in Figure 3(a)
is analysed in the present study. The building erected at the ELSA included the three interior
moment-resisting frames along with the secondary beams and the transverse cross bracing in
order to optimize the experimental effort. Main geometric characteristics of the prototype structure
realizing 3D specimen and design loads [15] are summarized in Figures 3(a)–(c). The prototype
structure was designed according to Eurocode provisions for all static load combinations involving
gravity, wind, snow and live loads, and seismic combinations. Wind loading was not found to be
critical both at the serviceability and at the ultimate limit states.
3.2. Earthquake resistance
Seismic actions on the structure were determined using the EC8 lateral force method of analysis
assuming importance category III (ordinary buildings) [16]. In the design stage, the frame was
assumed to be constructed on rock in an active seismic region with a design ground acceleration,
ag , equal to 0.4g. The latter is representative of regions exposed to very high seismic risk in
Europe, such as Turkey and Greece [28, 29]. For low-rise structures the design base shear force
reads Fb = 1.0× Sd (T1 )W , where Sd (T1 ) is the design spectrum ordinate reduced by behaviour
factor q at the fundamental period of the structure T1 . W is the structure seismic weight. For
design, the period T1 was taken to be equal to 0.05× H 0.75 = 0.215 s (H = 7.0 m), according to
the simplified method proposed by EC8 Type 1 spectrum suitable for magnitude 5.5 or greater
earthquakes was used for the structure assuming a ground type A (rock) [29]. Figure 4 shows the
elastic and inelastic design spectra provided by EC8 [16]. For T1 = 0.215 s, Sd = 0.167g resulting
in a base shear force of 0.167 W. The seismic weight included the dead load plus a fraction of
the imposed live loads, i.e. 48% at storey 1 and 60% at storey 2, giving W1 = 816.54 kN and
W2 = 839.01 kN, respectively. The design was performed for the two outmost interior frames for
which the seismic load was increased by 15% to account for accidental torsion. The total design
base shear was obtained to be Fb = 276 kN. According to EC8 provisions [16], 67% of that load
was applied at the top level and the remaining at the first level, assuming a triangular profile for
lateral force path.
3.3. Capacity design strategy
A beam end-plate design was selected for the PS joints because of its popularity in Europe and
because its inelastic rotation capacity had been extensively demonstrated in past research [30].
Well-proportioned column panel zones exhibit stable hysteretic shear response with significant
strain hardening behaviour [4, 31]. The potential for ductile and robust performance of connections
with energy dissipation shared between the beam end-plate and the column panel zones was
demonstrated for PS joints with structural steel columns [4, 8].
The selected connection type is not prone to premature beam weld fracture associated with
large column web shear deformations as observed in full-strength welded beam connections [32].
Shifting of beam hinging to shear yielding in column panel zone in multi-storey structures with fullstrength connections can result in undesirable column hinging patterns and storey mechanisms, with
concentration of inelastic demand and larger P– effects capable of causing structural collapse.
Sharing the inelastic demand between the beam end-plate and the column panel zone mitigates
this behaviour.
A strict capacity design procedure must, however, be followed to achieve a well-balanced
contribution to the inelastic response of these two connection components. Elastic analysis under
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
A. BRACONI ET AL.
Spectral Acceleration (g)
1.0
0.8
0.6
0.4
0.2
0.0
0.0
0.5
(a)
1.0
1.5
Period (s)
0.6
0.4
a (g)
0.2
0
0
2
4
6
8
10
12
14
16
18
-0.2
-0.4
-0.6
(b)
Time (sec)
12
EC8 code spectrum
Artificial spectrum
10
2
Sa (m/sec )
8
6
4
2
0
0
0.5
1
(c)
1.5
2
2.5
3
3.5
4
Period (sec)
Figure 4. Seismic action: (a) Elastic and design EC8 spectra; (b) Artificial earthquake ground
motion selected for the PSD test programme; (c) spectrum compatibility of artificial earthquake
with 5% damped EC8 response spectrum.
the prescribed Eurocode 1 [33] load combinations assuming rigid connection properties has been
firstly performed and preliminary beam and column sizes were selected to meet both the prescribed
serviceability and ultimate limit states. Beam end-plates and web column zones are then designed
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
STEEL–CONCRETE COMPOSITE MOMENT-RESISTING STRUCTURE
to also resist the actions obtained from the same elastic analysis. Their strength should be kept close
so that yielding will develop simultaneously in both components. In order to ensure an effective
hierarchy of yielding under strong ground motion shaking, beams, columns, and components have
to be checked against design forces compared with those that lead to yield dissipative mechanisms
in the end-plates and column web panels. The expected capacity of these ductile components was
determined with assumed steel yield strength equal to 1.3 times the nominal value for the check
of the members or connection components against brittle failure mode, such a flexural failure of
the columns or compression failure of the concrete slab bearing against the column face. The
various connection components are illustrated in Figure 1(b) with the indication of the failure mode
(brittle or ductile) considered. For end-plate bolts, premature failure of the bolts prior toflexural
yielding of the end-plates was prevented, satisfying the following conditions: t0.36d f ub / f y
[4, 5], where d and f ub are the diameter and nominal tensile stress of bolts, and t and f y are the
thickness and the nominal yield strength of the end-plate, respectively.
Once the capacity check is completed for all members and connection components, the connection stiffness properties can be determined and the process must be repeated taking into consideration the connection flexibility. Since beam end-plates and column web panels play a key role on
the connection stiffness and drift limits often govern the member sizes in moment frames, these
parts should be proportioned at this stage such that code drift limits can be met without further
increase in member sizes. The design of the shear connectors and the final layout of reinforcing
steel in the concrete slab are completed at the end of the design process.
3.4. The prototype structure
The nominal properties for the various steel materials are provided in Table I. The design was
performed according to the strategy described above except that the beneficial effect of the connection flexibility on beam-to-column joints was not considered in the design process. The resulting
beam and column sizes are shown in Figure 3(c). Full shear connection was provided between
the IPE300 beams and the concrete slab, as suggested by EC8 to avoid any interference between
low-cycle fatigue of connections and inelastic phenomena in joints. Two 19 mm studs were needed
at every deck flute to meet this requirement, as shown in Figure 3(c). Secondary beams were made
of IPE240 shapes connected to the columns through flexible thin plates. No shear connection was
provided between these beams and the concrete slab. Although these two details may not reflect
current construction practice, they were adopted to minimize the contribution of the secondary
beams to the frame lateral response and ease the analysis and interpretation of test results.
In the frame design, the columns were assumed to be fixed at their bases. For this structure,
column sizes were governed by code drift limitations and the selected cross sections are illustrated
in Figure 3(c). Longitudinal 12 mm and transversal 8 mm rebars were placed in the concrete portion
of the columns. At the base and near the beam-to-column joints, the stirrups were spaced at 50 mm
for both column types; see Figures 2(a) and (b). Elsewhere, the spacing was increased to 150 mm.
Tests by Elghazouli [34] and Takanashi and Elnashai [35] on similar partially encased composite
columns subjected to constant axial loading and cyclic rotational demand indicated that this column
design could undergo plastic rotations up to approximately 20 mrad prior to occurrence of local
buckling of the steel flanges and crushing of the concrete, thus enabling the development of the
full plastic mechanism intended in design with plastic hinges forming at column bases.
Flexible 15 mm thick end-plates were selected for all joints, and column stiffeners were used
at both beam flange levels to fully exploit the shear strength and inelastic deformation capacity of
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
A. BRACONI ET AL.
Table I. Nominal and actual steel and concrete material properties.
Component
f y,nom f u,nom f u,nom / u,nom
fy
fu
u
f cm
(MPa) (MPa) f y,nom (%) (MPa) (MPa) f u / f y (%) f y / f y,nom f u / f u,nom (MPa)
Structural Steel S 235 J0
IPE300
Flange 235
Web
235
IPE240
Flange 235
Web
235
HEB280
Flange 235
Web
235
HEB260
Flange 235
Web
235
End-plates
235
Steel B450 C
Reinf. bars
450
360
360
360
360
360
360
360
360
360
1.53
1.53
1.53
1.53
1.53
1.53
1.53
1.53
1.53
28
28
28
28
28
28
28
28
28
313
370
315
347
300
341
341
406
383
480
489
448
454
430
450
449
486
543
1.53
1.32
1.42
1.31
1.43
1.32
1.31
1.20
1.42
30.7
35.6
31.0
32.6
37.1
34.5
35.7
31.8
31.5
1.33
1.57
1.34
1.48
1.28
1.45
1.45
1.73
1.63
1.33
1.36
1.24
1.26
1.19
1.25
1.25
1.35
1.51
> 7.5
537
608
1.13 9.11
1.19
<1.17
>1.00
1.11
—
—
1130
—
20
383
550
—
—
—
—
—
—
—
—
>518 >1.15
<608 <1.35
Structural high strength bolts
M10.9
900
1000
Bolts
Structural Steel S 355 J0
Anchor
355
—
bolts
Concrete C 25/30
Slab
—
—
Columns
—
—
—
—
1.44 29.8
—
—
—
—
—
1.13
1.08
—
—
—
—
—
33
37.6
the web panel zone, as illustrated in Figure 1(a). Under negative moment, part of the tension force
acting at the top beam flange level is resisted by the slab reinforcing steel and a flush end-plate
detail was adopted. Conversely, an extended end-plate solution was chosen at the beam bottom
flange to resist the positive bending moments. The transfer of the compression force acting in the
slab to the column was assumed to be ensured through the two mechanisms shown in Figure 2(c)
[16]. Mechanism 2 requires additional transverse slab rebars to form the tension tie resisting the
transverse components of the two compressive struts near the column face.
At the design stage, a preliminary verification of the rotational capacity of the joints under
monotonically increasing loading was performed using the component model method [20] and
available data published in the literature on the ultimate deformation capacity of the joint components [36]. In all cases, the connections were found to reach the EC8 minimum value of 35 mrad
without strength degradation [16]. Detail on this verification can be found in [4, 36]. In the AISC
seismic provisions [18], the nominal strength of the joints in composite PR moment frames must be
at least equal to 50% of plastic flexural strength of the connected steel beams (ignoring composite
action). For exterior joints of the prototype structure, the ratios of the joints to steel beam nominal
flexural strength were 1.10 and 0.81 for positive and negative moments, respectively. For interior
joints, the corresponding values were 1.24 and 1.10. Hence, the joints as designed essentially met
and exceed by far this AISC minimum connection strength requirement [18]. Values of composite
steel–concrete joint resistances around the steel beam bending resistance are a consequence of the
drift control and resistance requests for static load combinations at the design stage.
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2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
STEEL–CONCRETE COMPOSITE MOMENT-RESISTING STRUCTURE
4. PSD AND CYCLIC TEST STRUCTURE
4.1. Description and installation
The test structure constructed at the ELSA of JRC included three of the five moment-resting frames
of the designed prototype structure, as illustrated on the plan view of Figure 3(b). The structure had
the same dimensions except that the slab overhangs extended to 700 mm in the transverse direction.
The modification was necessary to provide at least the effective slab width dimensions that were
assumed in the calculation of composite beam flexural properties. This layout also allowed proper
development of the slab anchoring for exterior beam-to-column joints. As also illustrated in Figure
3(b), the floor slabs at both levels were thickened and widened over a width of 1.5 m to form a
strong and stiff horizontal girder for the transfer of the loads induced by the actuators located at
each side of the test structure.
The columns were prefabricated in single two-storey pieces with column base joints and beam
connection details. The concrete fill was put in place in the horizontal position in the laboratory
before the erection of the structure. The columns were supported on isolated reinforced concrete
pedestals anchored to the laboratory test floor as shown in Figure 2(d). The column base joints
were endowed with 40 mm thick extended end-plates connected to the foundation by means of six
anchor rods made with 32 mm threaded hooked Fe510 anchor rods [15]. A 150 mm long shear
stub made from a HEB140 profile was welded under each base plate to transfer horizontal shear
forces to the foundation. Base plate stiffeners were installed on each column side to improve joint
fixity as depicted in Figure 2(d). Figure 5(a) shows the structure after completion of the erection.
In the design of the prototype structure, columns were assumed to be fixed at their base with
flexural yielding developing in columns, above their bases. In the design of each test frame, the
base plate and the anchor rods were designed using forces associated with the nominal plastic
flexural capacity of the composite columns, without applying the 1.3 overstrength factor to take
advantage of the inherent ductility of the joint at the ultimate limit state. Therefore, the column
base joints were expected to contribute to the energy dissipation capacity of the test structure.
4.2. Gravity loading
The total weights of the deck-slab assembly and steel profiles of the test structure at Levels 1 and 2
were 454 and 415 kN, respectively. Prior to starting the PSD test programme, additional gravity
loads of 518 and 496 kN were, respectively, applied at the first and second storeys representing
the additional dead load and the reduced imposed live load. This was achieved by the placement
of tanks filled with water on the slab at Level 1, as illustrated in Figure 5(b), and the addition of
two 20 ton concrete blocks at the top level.
4.3. Lateral loading and instrumentation
Lateral forces and displacements at each level were imposed by two 1000 kN hydraulic actuators
mounted horizontally on each side of the structure. The actuators were attached to the 16 m tall reaction
wall of the ELSA facility. Figure 5(b) shows one of the actuators as well as the transfer girders built in the
specimen floor slabs. During the tests, the lateral loads and displacements applied by the actuators were
recorded on a continuous basis as they were utilized in PSD control loop, as described in the companion
paper [17]. Two of the three frames of the test structure were extensively instrumented [14, 15] in order
to capture simultaneously both the global behaviour and the distribution of damage localized in the
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
A. BRACONI ET AL.
Figure 5. 3D full-scale prototype: (a) complete test frame erected at the ELSA facility (reaction wall
behind) and (b) additional gravity loads, i.e. water tanks, at floor 1.
joints and columns. Strain gauges were installed on slab rebars. They were also extensively used in
the columns on the middle frame so that internal member forces could be determined. Displacement
transducers and inclinometers were mounted to monitor the deformation of joints for an interior joint.
Inclinometers were also used at the column bases.
4.4. Actual material properties
Elementary tests were conducted to measure actual mechanical properties of various materials used
in the fabrication of the test structure. Table I compares the nominal and mean actual properties for
the various steel components and shows the measured mean values of the strength at 28 days of
the concrete. Most steel parts had much higher yield strength compared with the nominal values,
especially for the structural steel.
EC8 specifies that the actual yield strength used in the construction be such that plastic hinges
location assumed in design is not modified [16]. In order to avoid storey mechanisms or premature
brittle failure owing to the scatter of material strength values, the capacity design criterion of
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2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
STEEL–CONCRETE COMPOSITE MOMENT-RESISTING STRUCTURE
members and components of PS joints was checked using the measured material properties and it
was found that the intended yielding mechanism could be maintained.
5. JOINT RESPONSE AND NUMERICAL MODEL
5.1. Behaviour of PS joints
Prior to performing the PSD test programme, full-scale subassemblage monotonic and quasi-static
cyclic tests were performed at the University of Pisa on interior and exterior beam-to-column joint
specimens to validate design assumptions and to obtain data required to develop a numerical model
capable of predicting the behaviour of the test structure [4]. Typical test results are illustrated in
Figure 6. Very satisfactory inelastic responses with deformation levels exceeding EC8 minimum
requirements for DCH systems were obtained in all cases as extensively described in [14, 36].
Extensive yielding developed in the beam end-plates and the non-composite column web panel
zones is consistent with design assumptions. The behaviour of joints was also characterized by
inelastic straining of the slab reinforcement and crushing of the concrete slab against the column
[36]. Concrete crushing resulted in the sudden drop in resistance, which can be observed in Figure 6
on the sagging moment side. Close examination of the test results indicated that failure of the
concrete occurred because Mechanism 2 assumed for the transfer of part of the slab compression
force to the column could not be activated in the proposed joints, owing to a lack of continuity
between the concrete of the slab and the concrete fill of the columns [4]. This behaviour, however,
had small impact on the cyclic inelastic response capacity of the joints and the design was not
modified for the PSD test structure. In later applications of these joints, a shear transfer system
was conceived at the interface between the slab and column concrete materials so that Mechanism
2 could be fully exploited [37].
5.2. Numerical model
The PSD and cyclic quasi-static test programme was established based on the results of non-linear
static and dynamic time-history analyses of the test structure, allowing the selection of earthquake
ground motion excitations with acceleration levels suitable for each limit state. A 2D model
(Figure 7(a)) of one of the three moment-resisting frames of the test structure was developed using
the IDARC2D computer program [38]. The floor elevations were set at the centre of gravity of the
composite beams. The rotational behaviour of beam-to-column joints and column base joints was
simulated using hysteretic rotational springs located at the ends of rigid or beam–column elements.
The web panel shear distortion was modelled by four rigid bars connected by pins and rotational
springs. The behaviour of the frame sections and the rotational springs was simulated by means of
a smooth hysteretic model developed by Sivaselvan and Reinhorn [39]. Columns and beams were
introduced by using frame elements with spread plasticity and member properties were determined
from the measured material properties. The effective width of the concrete slab was established
according to EC8 [16] recommendations and the experimental response of beam-to-column joints.
A detailed non-linear cross-section fibre analysis was performed to establish the moment–curvature
response of columns. In these calculations, confinement effect on concrete properties based on
the model by Mander et al. [40] was considered for part of the concrete section. The hysteretic
parameters for beam-to-column joints were calibrated on the results of the subassemblage tests
performed at the University of Pisa. Steel components used in the aforementioned joint specimens
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
A. BRACONI ET AL.
Total Reaction [kN]
100
Monotonic Test
80
Cyclic Test
60
40
20
0
-20
-40
-60
-80
-100
-200
-100
0
100
(a)
Displacement [mm]
200
500
Experimental Data
400
M (kNm)
Numerical Simulation
-50
300
200
100
0
-40
-30
-20
-10
-100
0
10
20
30
40
50
-200
-300
-400
-500
(b)
Rotation (mrad)
300
Experimental Data
Numerical Simulation
250
200
M (kNm)
150
100
50
0
-20
-15
-10
-5
0
5
10
15
20
-100
-150
(c)
Rotation (mrad)
Figure 6. Response of composite joint specimens: (a) moment-rotation response from
full-scale subassemblage monotonic and quasi-static cyclic tests on an exterior joint specimen;
(b) hysteretic response of column web panel in shear and calibration of rotational spring
element representing its behaviour; and (c) hysteretic response of beam-to-column connection
and calibration of rotational spring element representing its behaviour.
were fabricated from the same material as the test structure, which enabled direct calibration of the
model with joint test results. Figures 6(b) and (c) show that a good correlation was obtained from
this calibration process, in particular, the connection and shear panel responses were calibrated
in order to fit mainly the hysteretic range and to reasonably reproduce their elastic behaviour.
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
STEEL–CONCRETE COMPOSITE MOMENT-RESISTING STRUCTURE
Figure 7. Numerical model of a test frame: (a) 2D finite element model; (b) model of the
base joint; and (c) lateral load–roof lateral displacement prediction of the test structure
under non-linear incremental static analysis.
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
A. BRACONI ET AL.
The rotational spring used in correspondence to the column base plate account represents the
inelastic behaviour of the anchor rods in tension and the grout in compression as depicted in Figure
7(b). As described also in the companion paper [17], the base plate representation was modified
to better capture the observed response of the as-built base plates.
5.3. Column bases
Experimental tests on column bases were carried out on full-scale subassemblages at the Laboratory
of the University of Naples Federico II, as depicted in Figures 8(a) and (b). This activity was part
of a collaborative research with the University of Sannio and later on with the University of Molise.
The objective of the experimental programme was to assess the rotation capacity of column bases
to be used in the test structure at the ELSA [27] and also to compare the performance of traditional
base plate connections with different layouts, inspired by those used for precast concrete frames,
i.e. socket-type connections.
The values of axial load (N ) used in the tests were equal to 170 and 330 kN, respectively.
The concrete class is C25/30, rebars are B450C class, and the structural profile is made by
S235 steel. These values correspond to the minimum and maximum axial loads relative to the
design load combinations of the full-scale composite framed building. Test results of the partially
encased composite columns were evaluated in terms of both local (moment–curvature M– and
moment–base rotation M–) and global (lateral load–top displacement, F–) response parameters.
For instance, the capacity curves of the specimens with HEB260 profiles, subjected to axial load
N = 330 kN, equipped with traditional and innovative base connections, are provided in Figure 8(c).
It is observed that the traditional connection layout exhibits lateral strength and stiffness higher
than the socket-type connection owing to the steel stiffeners used at the base of the column
and to the overstrength for seismic design [41]. Conversely, the ultimate deformation capacity of
the socket-type connection is about 75% higher than the traditional counterpart, i.e. 0.05 versus
0.09 rad. Furthermore, the failure mode of the specimen with steel end-plate is related to anchorage
bolt fracture, while in the case of the socket type the failure mechanism is related to plastic
deformation of flanges. However, in both cases the requirement of a minimum plastic rotation of
35 mrad provided by Eurocode 8 [16] is fulfilled, classifying both solutions as efficient for the
seismic design.
In particular, the composite partially encased columns with traditional connection yield for
a lateral load of 310 kN, which corresponds to a lateral drift of 26 mm (d/h ∼ 1.65%) under
monotonic regime. In both specimens, loaded with N = 170 and 330 kN, respectively, the column
strength and the energy dissipation do not exhibit significant decrease for drift d/h ∼ 5–6%. The
thick steel plate and the stiffeners used in the column base ensure that the end section of the
column remains plane. Under load reversal, the crushed concrete and the inelastic deformations
in the anchorages, both at the column base, could endanger the global lateral stiffness of the
composite column. For this reason, bond-slip phenomena between yielded steel of anchorage bars
and concrete and the degrading effects, especially at large drifts, that could reduce significantly the
energy dissipation capacity of the member have been examined more in detail. The behaviour of
anchorage bars embedded in the concrete foundation was performed by purposely specific pull-out
tests, as shown in Figures 9(a) and (b), respectively. These latter tests leaded to the definition of
adequate constitutive relationships (force-slip law) for this component of the base joint that was
used to refine the numerical model of the frame (as depicted in Figure 7(b)) for the numerical
simulations of the experimental response [27]. Moreover, the performance in terms of force and
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
STEEL–CONCRETE COMPOSITE MOMENT-RESISTING STRUCTURE
Figure 8. Test on composite column bases: (a) layout of the test set-up; (b) location of displacement
transducers; and (c) comparison between tests for HEB260 with different base solutions at N = 330 kN.
ductility (Figure 9) of the anchor bars coupled with the deformative capacity showed by the
traditional column base connection (Figure 8) assures a sufficient plastic rotation capacity to the
solution adopted for the 3D full-scale test frame.
5.4. Modal and pushover analyses
The periods in the first two modes of vibration of the test structure were obtained from modal
analysis: 0.43 and 0.13 s, respectively. It is noted that the fundamental period is nearly two times
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
A. BRACONI ET AL.
(a)
400
Test #1
Test #2
Force (kN)
300
200
100
0
0
(b)
5
10
15
Slip (mm)
20
25
Figure 9. Pull-out test for anchorage bolts for traditional base connections:
(a) set-up of the test and (b) test results.
longer than the value of T1 used in design (0.22 s), due to the simplified formula proposed by
EC8 for the fundamental period estimation that overestimates the lateral stiffness of the building.
Incremental static (pushover) analysis was then carried out to assess the as-built lateral capacity of
the test structure and to evaluate overstrength phenomena, using the model calibrated on the cyclic
tests made on sub-structures (joints and column bases); the adopted material resistances were those
coming from qualification tests executed on profile specimens and concrete. The model was also
used for more extensive static and dynamic studies performed to provide a more detailed assessment
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
STEEL–CONCRETE COMPOSITE MOMENT-RESISTING STRUCTURE
of the expected frame seismic performance [17] on the basis also of executed PSD programme.
In this first analysis, the gravity loading corresponding to Eurocode 1 [33] load combinations
was first applied to the structure and two lateral load conditions were considered, according to
EC8 suggestions [16] for structural assessment: uniform pattern and first mode shape distribution.
Computed normalized lateral load, V /W , versus normalized lateral roof displacement, /H , is
plotted in Figure 7(c). The two curves are quite similar, though they reflect the influence of larger
lateral forces applied at the top of the building with the first mode shape distribution. In both cases,
the first significant deviation from a linear response occurs at V /W = 0.5, which is approximately
2.5 times the EC8 design seismic loads, which corresponds to V /W = 0.206.
The first yielding is related to two factors: (i) the selection of members and drift limitations
under the serviceability load conditions and (ii) the scatter between nominal and actual yielding
stresses of steel profiles. These facts are confirmed by results of elastic analyses carried out
to assess the first yielding of the structure: design strengths of materials lead to V /W = 0.29;
the average yielding stresses for steel and compressive strength for concrete components lead to
a V /W = 0.50, calculated from pushover curves (Figure 7(c)). The latter show that the lateral
capacity near collapse reaches on average V /W = 1.2; hence, it can be argued that strain hardening
contributes significantly to the response of the structure and ensures a relevant overstrength ratio,
quite larger than the valued assumed for design (approximately estimated as u /1 2). However,
strain hardening and scattering of material properties are not the only causes of overstrength
phenomena. An interesting aspect is related to the drift limit checks, EC8 [16], with reference to
joint flexibility. From Figure 7(c), the computed drift angle under the design seismic load is 0.20%.
In EC8 [16], this deformation must be amplified by the q factor (6.0) and multiplied by the return
period reduction factor = 0.4 (Importance Classes III and IV), thus providing a value of ∼ 0.50%.
This value is equal to the EC8 limit for structures including brittle non-structural components.
Such a drift imposes a demanding checking of the limit state of damage (serviceability limit state,
SLS) that can lead to members oversized for dissipation purposes at limit state of safety (ultimate
limit state, ULS). A more harmonized definition of the suggested drift limits associated with the
expected performance as indicated in [16] could lead to an effective optimization of the member
sizes for the two considered limit states.
6. PSD AND CYCLIC TEST PROGRAMMES
6.1. Ground motion selection and dynamic analysis
In order to evaluate the response of the test structure under code compatible seismic demand at
all natural frequencies, a suite of artificial accelerograms were generated to match the EC8 Type 1
elastic response spectrum [42], i.e. Se . The ground motions were generated using the technique
described in Clough and Penzien [43]. The accelerogram that was selected for the PSD tests is
illustrated in Figure 4(b) and its 5% damped response spectrum is compared with the code spectrum
in Figure 4(c). This time history was selected, between all those generated, because it caused the
highest level of damage in beam-to-column joints and limited damage induced in columns of the
IDARC model. The ground motion time history has a peak ground acceleration (pga) of 0.46g.
It is characterized by 10 s strong motion duration, as prescribed in EC8 [16], with rise and decay
periods of 2.5 and 5.0 s, respectively. Trial non-linear time step analyses of the test structure were
performed using the IDARC frame model [38] to establish ground amplitudes required to reach
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2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
A. BRACONI ET AL.
a set of predetermined limit states. The analyses were performed using the trapezoidal rule in
the implicit -Newmark time-stepping scheme [44], with a single correction. Time steps were set
equal to 0.001 and 0.0001 s used for the elastic and inelastic analyses, respectively; the Rayleigh
damping was set to 5% of critical damping in the first two modes of vibration. More details about
the whole test programme are reported in the following section, while main results coming from
numerical simulations and tests are reported in the companion paper [17].
6.2. PSD and cyclic test programme
In accordance with the performance-based earthquake engineering approach, it was decided to
carry out a series of four PSD tests at increasing ground motion amplitudes to examine the response
of the structure corresponding to various limit states or anticipated performance levels. The chosen
test programme is reported hereafter with the performance objective (PO) foreseen:
• PSD test 1—pga set equal to 0.10g; PO: elastic response;
• PSD test 2—pga set equal to 0.25g; PO: no structural damage;
• PSD test 3—pga set equal to 1.40g; PO: joint plastic rotation near 35 mrad with less than
20% of strength degradation;
• PSD test 4—pga set equal to 1.80g; PO: joint plastic rotation beyond 35 mrad.
A final cyclic quasi-static test with stepwise increasing large amplitudes was then added to
induce severe damage to the structure and allow failure mechanisms of the structure components to
be examined. The analyses indicated a nearly elastic structural response without concrete cracking
under the ground motion time history scaled to 0.10g pga. A PSD test was therefore proposed at
this amplitude to evaluate the dynamic elastic uncracked properties of the structure, including its
natural frequencies, associated mode shapes, and damping levels. This test was also carried out
to verify the adequacy of both the test set-up and PSD algorithm without damaging the structure.
An SLS was established at a 0.25g pga, a ground motion level that brings the structure near first
significant yielding with peak storey drift angles reaching approximately 1%. Such a ground motion
amplitude is associated with a probability of exceedance of 10% in 10 years in high seismic sites.
In order to reach rotational demand in joints up to or close to the EC8 requirement of 35 mrad [16],
the accelerogram amplitude had to be increased such that its pga reached 1.40g. This corresponds
to 3.5 times the design acceleration level of 0.40g, a difference consistent with the results of the
static incremental analysis. The PSD Test No. 3 performed at this amplitude therefore aims at
examining the global performance of the structure at the ULS, with specific interest in the response
of the PS beam-to-column joints at levels corresponding to the EC8 prescribed rotation limit [16].
The anticipated peak storey drift angles under this ground motion level are equal to 2.5%, thus 2.5
times the values expected at the SLS. Figure 10(a) shows the predicted top storey displacement at
this earthquake level. The PSD Test No. 4 is performed to observe the behaviour of the frame for
total rotation just beyond the 35 mrad inelastic rotation requirement of EC8 and a pga value of 1.8g
was selected to impose such a demand to the structure. Under such a ground motion amplitude,
the analysis predicted plastic rotation of up to 40 mrad in the joints and interstorey drift angles
of 4.5%, thus close to the 0.04 rad specified in the AISC seismic provisions for composite PR
moment-resisting frames. The final quasi-static cyclic test was developed according to the ECCS 45
procedure [45] with stepwise increasing amplitude cyclic displacements in order to induce a severe
amount of damage in beam-to-column joints, column base joints, and columns in a controlled
and systematic manner. The imposed roof displacement history is illustrated in Figure 10(b).
Copyright q
2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
STEEL–CONCRETE COMPOSITE MOMENT-RESISTING STRUCTURE
400
2
1.5
1
0.5
0
-0.5 0
-1
-1.5
-2
-2.5
-3
Top Storey
300
200
100
2
4
6
8
10
12
14
16
0
18
-100
-200
-300
(a)
(b)
Time (sec)
-400
3
0.8
Experimental Data
Numerical Simulation
0.6
Experimental Data
Numerical Simulation
2
0.4
1
0.2
0
0
0
0
2.5
5
7.5
10
12.5
-115
2.5
5
7.5
10
12.5
-1
15
17.5
-1
17.5
-0.2
-2
-0.4
-3
-0.6
-4
-0.8
(c)
Time (sec)
(d)
Time (sec)
Figure 10. Predicted response of the roof displacement: (a) time history under ground motion scaled at
1.4g pga and (b) history developed for the cyclic quasi-static test. Measured and predicted top storey
displacement time histories: PSD Test No. 2 pga=0.25g(c); PSD Test No. 3 pga=1.40g(d).
A total of six displacement increments with two cycles per increment were applied for a total of
12 cycles. The maximum displacement amplitude was equal to 300 mm, with predicted interstorey
drift angles of 4.6%. During this test, the first to the second floor lateral load ratio is maintained
equal to 0.97. This ratio was determined from modal shapes obtained in the PSD Test No. 4.
7. CONCLUSIONS
A comprehensive design procedure, applied to a prototype structure, was proposed for steel–
concrete composite moment-resisting structure endowed with PS beam-to-column joints designed
to dissipate seismic energy through bending of the beam end-plates and shear yielding of column
web panel zones. The performance of joints was verified through subassemblage quasi-static cyclic
tests. The results were used to calibrate a numerical model developed to predict the seismic
behaviour of a full-scale structure specimen to be used in a PSD test programme. The frame was
found to exhibit significant extra lateral capacity compared with the value expected in design. The
results of the analysis and the calibrated numerical model were used to develop a comprehensive
multi-level PSD programme targeted to analyse the response of the structure at various limit states
or anticipated performance levels. A final quasi-static cyclic test programme was also included in
the test programme to induce severe damage to the structure components. The calibrated model of
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2008 John Wiley & Sons, Ltd.
Earthquake Engng Struct. Dyn. (2008)
DOI: 10.1002/eqe
A. BRACONI ET AL.
the test structures adopted for the PSD test programme definition also demonstrated good agreement
with the experimental results (Figures 10(c) and (d)). The results coming from 3D full-scale test
are further discussed in the companion paper [17].
ACKNOWLEDGEMENTS
The results presented in this work were obtained in the framework of the ECOLEADER HPR-CT-199900059 and the ECSC 7210-PR-250 European research projects, for which the authors are grateful. The
last author collaborated to this study as a Visiting Scientist at the Joint Research Centre at Ispra during
a sabbatical leave from Ecole Polytechnique of Montreal (Contract No. 19851-2002-09 P1VS3 ISP IT).
The financial support from both institutions and the Natural Sciences and Engineering Research Council
of Canada is acknowledged. Nevertheless, opinions expressed in this paper are those of the authors and
do not necessarily reflect those of the sponsors.
REFERENCES
1. Leon RT, Hoffman JJ, Staeger T. Partially Restrained Composite Connections. Steel Design Guide Series, vol.
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